[fusion_builder_container hundred_percent="no" equal_height_columns="no" hide_on_mobile="small-visibility,medium-visibility,large-visibility" background_position="center center" background_repeat="no-repeat" fade="no" background_parallax="none" enable_mobile="no" parallax_speed="0.3" video_aspect_ratio="16:9" video_loop="yes" video_mute="yes" overlay_opacity="0.5" border_style="solid" padding_top="20px" padding_bottom="20px"][fusion_builder_row][fusion_builder_column type="1_1" layout="1_1" spacing="" center_content="no" hover_type="none" link="" min_height="" hide_on_mobile="small-visibility,medium-visibility,large-visibility" class="" id="" background_color="" background_image="" background_position="left top" background_repeat="no-repeat" border_size="0" border_color="" border_style="solid" border_position="all" padding="" dimension_margin="" animation_type="" animation_direction="left" animation_speed="0.3" animation_offset="" last="no"][fusion_text] QUESTION: (SUMMER2016) I am proportioning a mono-symmetric I-girder whose web-slenderness, h/w, marginally satisfies the Class 3 limit for W-shapes in pure bending in accordance with Table 2 of S16-09. Does this limit apply to mono-symmetric sections? ANSWER: No, the limits for I-sections are provided in Table 2 of S16-09 for sections with equal flanges. CSA Standard S6, Canadian Highway Bridge Design Code, in Clause 10.10.3.1, addresses Class 3 web-limits for mono-symmetric I-sections. In this clause, the value of h is replaced by 2d c , where d c is the distance from the neutral axis to the compressive extreme fibre (see Figure). [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (2015) When using the direct method in accordance with Clause 9.2.6.2 of CSA S16-14, I was surprised to find that, for a given column force, C f , the bracing force, P b , increases as the number of braces is increased from 1 to 2. Does this make sense? ANSWER: Intuitively, it would seem that adding braces to stabilize a compression member should reduce the brace force per brace, as there are more braces to “share” the stabilizing forces. But in fact, the opposite is true. It should be noted that the brace forces in two adjacent braces do not act in the same direction but the opposite is true (Figure 1). Given a column force, C f , and in compliance with a maximum permitted out-of-plumbness, ∆ 0 /L, the bracing force, P b , is directly proportional to the factor β, which takes on values of 2 and 3 for 1 and 2 equally spaced braces respectively. In this case, the brace force, P b , increases by 50%. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SUMMER2014) I am designing a building structure consisting of a simple gravity frame and a perimeter rigid frame, which serves as the lateral-force resisting system. There is no braced bay or shear wall. Several gravity columns are subjected to significant bending about the strong axis due to a large connection eccentricity. Should these wide-flange columns be designed as beam-columns of an unbraced frame? How are the values for U 1x determined? ANSWER: No, this is a braced-frame situation. Despite the absence of braced bent, the gravity columns are ‘leaners’, i.e. they do not participate as primary lateral-force resisting members. The U 1x values can be calculated in accordance with Clause 13.8.4 of CSA S16 but the U 1x values for determination of cross-sectional strength and lateral-torsional buckling strength must not be less than 1.0. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SPRING2014) I heard about the notional load requirement in the design of building structures but I cannot locate the notional load provision in the building code. Where do I find them? ANSWER: You will find the provision for notional loads in CSA Standard S16-09, Design of Steel Structures. The Standard permits the second order effects due to gravity loads acting on the displaced structure under horizontal loads to be accounted for by using a P-delta analysis. In addition, the effects due to out-of-plumbness and partial yielding may be approximated by a set of horizontal loads, which are referred to as notional loads. The notional load to be applied at each level in addition to any other horizontal load is taken as 0.5 per cent of the concurrent gravity load acting on that level. Alternatively, a rigorous second-order analysis that accounts for both geometric nonlinearity, including out-of-plumbness, and partial yielding may be used. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (FALL2013) When checking flexural buckling of a channel section under axial load, what radius of gyration should be used to calculate F ex and F ey ? ANSWER: In S16-09 Clause 13.3.2, the elastic buckling stresses are given by: For singly-symmetric sections such as channels, the same clause specifies that the y-axis is taken as the axis of symmetry . But when using the tables of properties and dimensions for channels in Part 6 of the Handbook of Steel Construction , the x-axis is defined as the axis of symmetry. Therefore, F ex should be calculated using the radius of gyration r y as given in the Handbook tables, and likewise F ey should be calculated using r x . [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (FALL2012) How do I determine the elastic bending resistance of a cantilever plate subject to lateral-torsional buckling? ANSWER: For a fix-ended plate subject to bending about its strong axis but laterally unbraced, the Guide to Stability Design Criteria for Metal Structures, 6 th Edition (R.D. Ziemian, John Wiley & Sons, 2010) gives expressions for the elastic buckling moment (M u ), depending on the height of load application. For example, when subjected to a point load at its tip (see Figure): a) For top surface loading: b) For shear centre loading: where E and G are the elastic and shear modulii respectively, I y is the minor-axis moment of inertia, J the St. Venant torsional constant, “d” the plate depth, and L c the cantilever length. NOTE: In Advantage Steel #44, this column referenced the expressions for the elastic lateral-torsional buckling moment of cantilevers provided in the Guide to Stability Design Criteria for Metal Structures, 6th Edition. In comparison with recent studies using finite element analyses, the expression “Mc = 1.5GJ/d” gives unconservative values for plates (rectangular section) and long cantilevers of I-sections prone to lateral-torsional buckling. It should not be used for plate cantilevers significantly longer than twice their depth. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SUMMER2012) In the design of continuous beams and Gerber beams can I assume the inflection points are laterally braced against lateral-torsional buckling? ANSWER: One must not confuse the inflection points in the vertical bending moment diagram with the inflection points in the laterally buckled shape. In general, the buckled shape is not known at the design stage. Since the inflection points in the bending moment diagram generally do not coincide with those of the buckled shape they should not be taken as laterally braced unless they are braced. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SUMMER2012) I have used the effective-length method to design columns in sway-permitted frames as well as those in sway-prevented frames. But I cannot find the effective-length nomograph for sway-permitted frames in CSA Standard S16. Is the effective-length method still valid? ANSWER: The effective column length method attempts to approximate the elastic critical column load for a regular frame that is without primary moments and consists of identical beams in each level and identical columns (see Figure 2). This idealized frame model does not account for any second-order effects in the beams at all. Moreover, in comparison with modern analysis methods that account for second-order effects etc., it usually fails to provide accurate results for real frames. Since the introduction of S16.1-M89, the Standard has abandoned the effective length design method for sway-permitted frames. Accordingly, the nomograph for sway-permitted frames has since been excluded. When an elastic analysis is used, the current standard, S16-09, requires the application of a second-order analysis that directly accounts for sway effects. Alternatively, P-delta effects are included using the amplification factor, U 2 . In addition, notional loads are applied to account for the effects of partial yielding and initial out-of-plumbness. Note: When P-delta effects and notional loads are accounted for, the use of an effective column length for sway-prevented case (K ≤ 1) is permitted for consideration of lateral-torsional buckling. However, it is assumed in the effective length method that all members remain elastic prior to buckling. Hence the strong-column-weak-beam requirements for non-conventional construction, where applicable, may render its use inappropriate. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SPRING2012) When I use the amplifier, U 2 , to account for P-Δ effects in accordance with S16, should I apply U 2 to amplify the notional loads as well? ANSWER: Yes, notional loads should also be amplified when U 2 is used to account for P-Δ effects. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: Part A: (FALL2011) How do I calculate the axial compressive resistance of a member subject to elastic local buckling? ANSWER: Two methods are provided in CSA Standard S16-09, the effective area method and the effective yield stress method. Effective area method Engineers are generally familiar with the concept of effective area when designing columns subject to elastic local buckling. Such sections are expected to undergo local buckling before reaching the yield load in axial compression, AF y . They are designed in accordance with CSA S16-09 Clause 13.3.5 (a) whenever the width-to-thickness ratio of the flanges or web exceeds the limits given in Table 1 of S16-09. When calculating the axial compressive resistance, a portion or portions of the cross-section is considered ineffective and is therefore omitted. Considering a wide-flange section, for example, the effective cross-sectional area, A e , is computed as follows: If the flanges exceed the maximum width-to-thickness ratio of Table 1, the area of the tips (shaded parts in the figure shown) is removed, such that the remaining effective flange width, b e , meets the maximum ratio; similarly, the effective web depth is taken as h e as shown in the figure. The effective portions of the flanges and web together make up the effective area, A e . Effective yield stress method Perhaps less familiar is the effective yield stress method, which S16-09 also permits for calculating the axial compressive resistance. According to this concept first introduced in S16-01, the cross-sectional area remains intact, but the yield stress is reduced to account for local buckling. The effective yield stress, F ye , is taken as the reduced yield stress determined from the width-to-thickness ratio meeting the limit in Table 1. If both the flanges and the web are subject to elastic local buckling, two separate effective yield stresses are calculated. For simplicity, the member resistance is based on the lower of the two values. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: Part B: (FALL2011) Do the methods provided in CSA S16-09 give the same answer? ANSWER: ANSWER: No, the effective area method and the effective yield stress method, in general, do not give the same answer. An example is presented below to illustrate both methods. Consider a laterally supported W360x72 column (L = 0) made of ASTM A992 steel. The cross-sectional area is A = 9100 mm 2 and the specified yield stress, F y , = 345 MPa. The factored axial compressive resistance will be determined on the basis of (1) effective area and (2) effective yield stress. (1) Effective area method Check the width-to-thickness ratios of the flanges and web : <img class="alignnone size-full wp-image-14698" src="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-1.jpg" sizes="(max-width: 249px) 100vw, 249px" srcset="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-1-120x29.jpg 120w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-1-200x49.jpg 200w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-1.jpg 249w" alt="" width="249" height="61" /> The flanges are not subject to local buckling. <img class="alignnone size-full wp-image-14699" src="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stabilty-equation-part-b-2.jpg" sizes="(max-width: 456px) 100vw, 456px" srcset="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stabilty-equation-part-b-2-120x17.jpg 120w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stabilty-equation-part-b-2-200x28.jpg 200w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stabilty-equation-part-b-2-300x42.jpg 300w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stabilty-equation-part-b-2-400x56.jpg 400w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stabilty-equation-part-b-2.jpg 456w" alt="" width="456" height="64" /> The web is subject to elastic local buckling. The effective web depth is given by: <img class="alignnone size-full wp-image-14700" src="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-3.jpg" sizes="(max-width: 170px) 100vw, 170px" srcset="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-3-120x42.jpg 120w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-3.jpg 170w" alt="" width="170" height="59" /> The effective area is: <img class="alignnone size-full wp-image-14701" src="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-6.jpg" sizes="(max-width: 242px) 100vw, 242px" srcset="https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-6-120x15.jpg 120w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-6-200x26.jpg 200w, https://cisc-icca.ca/ciscwp/wp-content/uploads/2017/02/stability-equation-part-b-6.jpg 242w" alt="" width="242" height="31" /> And the compressive resistance is: (2) Effective yield stress method The effective yield stress is based on the maximum width-to-thickness ratio of the web: And the compressive resistance is: Although only the web is subject to local buckling, the entire cross-section is affected by the reduced yield stress. For this reason, the effective area method results in a greater resistance (for a laterally braced member) than the effective yield stress method in this particular example. However, the effective yield stress method usually gives a larger resistance for slender members.For both methods, the elastic buckling stress, F e , is determined using gross section properties. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: Part C: (FALL2011) Which method is used to calculate the values for sections subject to elastic local buckling in the Column Tables and Angle Strut Tables in the CISC Handbook? ANSWER: Both methods are used to calculate the factored compressive resistances, C r , for W-columns subject to elastic local buckling and the tabulated values are the larger of the two. Only the effective area method is used to calculate the C r values for other columns and struts. Angle sections that exceed the maximum b-to-t limit in Table 1 are excluded from the star-shaped angle strut tables. [/fusion_text][/fusion_builder_column][/fusion_builder_row][/fusion_builder_container]

[fusion_builder_container hundred_percent="no" equal_height_columns="no" hide_on_mobile="small-visibility,medium-visibility,large-visibility" background_position="center center" background_repeat="no-repeat" fade="no" background_parallax="none" enable_mobile="no" parallax_speed="0.3" video_aspect_ratio="16:9" video_loop="yes" video_mute="yes" overlay_opacity="0.5" border_style="solid" padding_top="20px" padding_bottom="20px"][fusion_builder_row][fusion_builder_column type="1_1" layout="1_1" spacing="" center_content="no" hover_type="none" link="" min_height="" hide_on_mobile="small-visibility,medium-visibility,large-visibility" class="" id="" background_color="" background_image="" background_position="left top" background_repeat="no-repeat" border_size="0" border_color="" border_style="solid" border_position="all" padding="" dimension_margin="" animation_type="" animation_direction="left" animation_speed="0.3" animation_offset="" last="no"][fusion_text] QUESTION: (2015) What are the most common grades for structural steel shapes and sections used in building construction? ANSWER: The main contributing factors are: a) their suitability for the intended applications as recognized by codes and standards, and b) availability. A summary for common structural steel grades used for building construction is shown in the table below: It should be noted that an A500 HSS is not an exact substitution for its G40.21 350W counterpart having the same nominal size designation, mainly due to the less stringent ASTM A500 under-mass tolerance and, in some cases, lower tensile strength properties. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (FALL2015) I am involved with the design of an unenclosed industrial structure for which neither the building code nor the bridge code applies. Should I specify notch-tough steel? Are notch-tough W-shapes available? How do I determine the appropriate level of notch-toughness? ANSWER: Brittle fracture is a complex subject. In order to limit the probability of its occurrence to an acceptable level, one should account for the consequence of failure together with many influential factors as described in Annex L of CSA S16-14. Notch toughness of steel is only one of these factors. Often, fracture control for structures other than buildings and bridges involves identifying the safe options then choosing the most viable solution. Hence the engineer(s) who is responsible for the design, construction specifications and supervision, QA, QC, recommendation for future inspections, etc. is in the best position to make this decision.With respect to steel grades and their availability, this is the current situation. Although common structural steels, such as CSA G40.21 Types W and A steels generally possess notch toughness that is superior to many steel products used for non-structural applications, they are not produced to meet specific impact testing requirements. Types WT and AT steels are. Purchasers of Types WT and AT steels must also specify the required notch-toughness category that establishes the Charpy V-notch test temperature and energy level. Similarly, purchasers of ASTM steel grades, such as A992, must specify the appropriate test temperature and energy level if they so desire. To our knowledge, a North American mill produces W-sections up to about 440 kg/m to notch-toughness requirements comparable to CSA 350WT Cat. 3. It should be noted that Charpy V-notch test requirements add cost and lead time. Therefore, they should not be specified indiscriminately. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (2014) Are ASTM A1085 products readily available? ANSWER: ASTM A1085 is a new standard, introduced in 2013. Atlas Tube Canada ULC has started taking orders in 2013. Time will tell if they will be readily available from service centres. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SUMMER2011) Why have Welded Wide Flange (WWF) shapes been commonly used in Canada? ANSWER: The introduction of CSA G40.12 in 1964 marked the beginning of an era for Canadian structural steel. This stronger (300 MPa specified yield) steel replaced ASTM A36 (248 MPa yield) steel as the basic grade for wide-flange shapes, etc. Algoma Steel, the Canadian wide-flange mill at that time, did not produce the full range of W-sections. As a result, the welded wide-flange (WWF) sections were introduced as alternatives for heavy rolled sections.As built-up shapes, the WWF cross-sections can be tailored for optimal efficiency; standard beam sections were compact sections or stockier and all columns were Class 3 or stockier. Because the plates were oxy-flame-cut in Algoma’s facility, their WWF sections were also qualified for the most favourable column design curve (of the three SSRC curves), whereas the least favourable curve applied to heavy A36 W-shapes.The above-mentioned advantages, coupled with the higher strength of the 300 MPa steel, had helped WWF sections to remain popular choices for heavy sections until ASTM A992 (345 MPa specified min. yield) emerged as the common North American steel grade for wide flange shapes about a decade ago. Until recently, WWF sections also benefited from the availability of notch-tough steel plates where required, while rolled shapes, supplied to meet certain notch-toughness requirements, were scarce. Rolled W-shapes, up to 440 kg/m in weight and meeting CSA Type T Category 3 notch-toughness requirements, are now available (subject to minimum tonnage order, etc.). All in all, the clear advantages that WWF shapes once offered have dwindled. [/fusion_text][fusion_separator style_type="single solid" hide_on_mobile="small-visibility,medium-visibility,large-visibility" top_margin="10px" bottom_margin="10px" border_size="2" alignment="center" /][fusion_text] QUESTION: (SUMMER2011) Essar Steel Algoma has closed their WWF production facility. What are the steel designer’s choices? ANSWER: For several years, another source of WWF sections had supplied the market in western Canada. Following Essar Steel Algoma’s exit, a large fabrication facility in eastern Canada has shown interest in WWF production. The future availability of WWF sections, or lack of, will be sorted out in the marketplace. It should be noted that WWF sections are essentially standardized welded built-up H-shapes (see answer to the question below). Hence they can be replaced with custom-designed welded built-up sections.Here are the current choices: a) North American mills produce ASTM A992 rolled W-shapes up to 1,100 mm in depth, and column sections weigh up to 1,086 kg/m. At current exchange rates, they ought to be very competitive; and b) Welded plate girders for sections deeper than 1,100 mm and where the need to camber heavy rolled girders renders them unsuitable. [/fusion_text][/fusion_builder_column][/fusion_builder_row][/fusion_builder_container]

FAQs Archive – Page 3 of 3

– Stability

QUESTION: (SUMMER2016) I am proportioning a mono-symmetric I-girder whose web-slenderness, h/w, marginally satisfies the Class 3 limit for W-shapes in pure bending in accordance with Table 2 of S16-09. Does this limit apply to mono-symmetric sections? ANSWER: No, the limits for I-sections are provided in Table 2 of S16-09 for sections with equal flanges. CSA [...]

CISCWebAdmin2017-10-11T14:35:01-04:00October 1st, 2017|

– Steel Properties & Grades

QUESTION: (SUMMER2016) For wide-flange sections used for building construction, should I specify CSA G40.21 Grade 350W or ASTM A992? ANSWER: ASTM A992/992M should be specified. It is the grade that North American wide-flange mills produced to and CSA Standard S16-14 (and S16-09) explicitly recognize. Introduced in the 1990s as a product with enhanced properties for [...]

CISCWebAdmin2017-10-03T15:40:38-04:00October 1st, 2017|

– Steel Supply

QUESTION: (2015) What are the most common grades for structural steel shapes and sections used in building construction? ANSWER: The main contributing factors are: a) their suitability for the intended applications as recognized by codes and standards, and b) availability. A summary for common structural steel grades used for building construction is shown in the [...]

CISCWebAdmin2017-10-03T15:40:53-04:00October 1st, 2017|

Industrial Structures & Platework – Industrial Structures

QUESTION: (FALL2015) I am involved with the design of an unenclosed industrial structure for which neither the building code nor the bridge code applies. Should I specify notch-tough steel? Are notch-tough W-shapes available? How do I determine the appropriate level of notch-toughness? ANSWER: Brittle fracture is a complex subject. In order to limit the probability [...]

CISCWebAdmin2017-10-03T15:45:16-04:00October 1st, 2017|

Seismic & Wind Effects – Seismic

QUESTION: (SUMMER2016) In National Building Code of Canada 2015, the period dependant Site Coefficient F(T), used for determination of seismic Design Spectral Accelerations, has replaced Site Coefficients Fa and Fv. CSA Standard S16-14, however, continues to reference Site Coefficients, Fa and Fv. Is S16-14 out of step with NBC 2015? ANSWER: CSA S16-14 is compatible [...]

CISCWebAdmin2017-10-03T15:44:25-04:00October 1st, 2017|

– Wind

QUESTION: (SPRING2012) The User’s Guide – NBC Structural Commentaries provides external peak pressure-gust coefficient values, in Figures I-8 to I-14, for design of cladding and secondary structural members. Should these values or those tabulated in Figure I-7 be used for the design of a primary roof girder in a low-rise building? ANSWER: The values tabulated [...]

CISCWebAdmin2017-10-03T15:44:37-04:00October 1st, 2017|